Report (2) ,
ecrtechnical Investigation and ;.
Seismic Site Hazard Investigation
Washington Square Mall - Phase .I
Tigard, Oregon
September 19, 2003
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Geotechnical Investigation and
Seismic Site Hazard Investigation
Washington Square Mall - Phase 1
ITigard, Oregon
September 19, 2003
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I SHANNON WILSON,INC.
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<. GEOTECHNICAL AND ENVIRONMENTAL CONSULTANTS
IAt Shannon& Wilson, our mission is to be a progressive, well-
managed professional consulting firm in the fields of engineering
1 I and applied earth sciences. Our goal is to perform our services
with the highest degree of professionalism with due consideration
Ito the best interests of the public,our clients,and our employees.
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MacerichSubmitCompany edTo:
16495 NE 74th Street
Redmond, WA 98052
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By:
I Shannon &Wilson, Inc.
2255 SW Canyon Road
Portland, Oregon 97201
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SHANNON FIWILSON.INC.
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TABLE OF CONTENTS
I Page
1 1.0 INTRODUCTION 1
2.0 SCOPE OF WORK 1
I3.0 SITE AND PROJECT DESCRIPTION 1
1 4.0 PREVIOUS GEOTECHNICAL INVESTIGATIONS 2
5.0 FIELD EXPLORATIONS AND LABORATORY TESTING 3
1 5.1 Field Explorations 3
5.2 Laboratory Testing 4
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6.0 GEOLOGY AND SEISMICITY 4
6.1 General Geology 4
6.2 Seismic Setting 5
1 6.3 Seismic Source Zones 5
6.4 Local Seismic Sources 7
6.5 Historical 7
I7.0 SITE CHARACTERIZATION 8
7.1 Subsurface Conditions 8
1 7.1.1 General 8
7.1.2 Mall Addition and North Parking Structure 8
7.1.3 South Parking Structure 9
1 7.2 Subsurface Profile 9
8.0 SEISMIC HAZARDS 10
1 8.1 Design Earthquakes 10
8.2 Ground Response 12
8.3 Seismic Hazards 12
I 8.3.1 General 12
8.3.2 Liquefaction 13
1 9.0
GEOTECHNICAL RECOMMENDATIONS 14
9.1 Earthwork 14
9.1.1 General Site Preparation 14
1 9.1.2 Excavations 14
9.1.3 Temporary Dewatering 14
9.1.4 Structural Fill 15
19.1.5 Slopes 15
9.2 Foundation Design 15
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rTABLE OF CONTENTS (cont.) - SHANNON&WILSON,INC.
I 9.2.1 General 15
9.2.2 Pile Foundations 16
9.2.2.1 Pile Types 16
I 9.2.2.2 Pile Capacity and Driving Criteria 16
9.2.2.3 Estimated Pile Lengths 17
9.2.2.4 Uplift Capacity of Piles 17
I 9.2.2.5 Lateral Pile Capacities 18
9.2.3 Spread Foundations 18
9.2.4 Floor Slabs 19
I 9.3 Embedded Walls 20
9.4 Pavement Design 21
9.4.1 Pavement Sections 21
I 9.4.2 Pavement Subgrade 21
9.4.3 Pavement Materials 21
I10.0 LIMITATIONS 22
REFERENCES CITED 24
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LIST OF TABLES
1 Table No.
I1 Summary of Borings
ILIST OF FIGURES
Figure No.
r1 Vicinity Map
2 Plan of Explorations
I 3 Soil Classification and Log Key
4 Log of Boring B-1
5 Log of Boring B-2
6 Log of Boring B-3
7 Log of Boring B-4
8 Log of Boring B-5
9 Log of Boring B-6
10 Log of Boring B-7
11 Log of Boring B-8
12 Log of Boring B-9
13 Log of Boring B-10
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14 Log of Boring B-11
15 Log of Boring B-12
16 Log of Boring B-13
17 Log of Boring B-14
I 18 Plasticity Chart
19 Grain Size Distribution
20 Consolidation Test
21 Footing Overexcavation Detail
22 Backfill &Drainage Details
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LIST OF APPENDICES
Appendix No.
A Logs of Pertinent Borings from Previous Geotechnical Investigations at
Washington Square
B Important Information About Your Geotechnical/Environmental Report
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GEOTECHNICAL INVESTIGATION AND
I SEISMIC SITE HAZARD INVESTIGATION
WASHINGTON SQUARE MALL- PHASE 1
TIGARD, OREGON
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I1.0 INTRODUCTION
This report presents the results of our geotechnical investigation and our seismic site hazard
I investigation and our comments and recommendations regarding earthwork, foundations,
pavement design, and seismic vulnerability. The purposes of the geotechnical investigation were
I to evaluate subsurface conditions at selected locations on the site and to assist with the design as it
relates to earthwork, foundations and pavements. The purpose of the seismic site hazard
investigation was to evaluate on a site specific basis the vulnerability of the structures to seismic-
', induced geologic hazards as required by Section 1804.3.2 of the 1998 Oregon Structural Specialty
Code based on the 1997 Uniform Building Code for essential facilities,hazardous facilities,major
Istructures and special occupancy structures. The parking structures included in project are
classified as major structures by the Code.
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2.0 SCOPE OF WORK
IOur scope of work included the drilling and sampling of 14 bore holes, completing geotechnical
laboratory tests on soil samples,performing engineering analyses, and preparing this report. Our
1 work was performed in accordance with the Macerich Company Professional Services
Agreement dated April 8, 2003, and Amendments 1 and 2 dated April 18, 2003 and August 5,
1 2003,respectively.
1 3.0 SITE AND PROJECT DESCRIPTION
The Washington Square Mall complex is located in a roughly triangular area bordered by Oregon
1 State Highway 217, SW Hall Boulevard, and SW Greenburg Road in Tigard, Oregon as shown
on the Vicinity Map, Figure 1. The original Mall construction consisted of an L-shaped structure
I with JC Penney, Meier& Frank and Sears as the anchor stores. Nordstrom was added later. The
Mall itself is a one level structure,but the anchor stores are two to three levels with the Sears
Istore having a basement.
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IThe proposed expansion of the Washington Square complex will include a one and two level
addition on the west side of the Mall in the paved parking area between Meier&Frank and
I Nordstrom. The new addition will be isolated from the existing Mall,but will be directly
adjacent to it. The addition will be about 235 feet by 545 feet in plan, and the finished floor of
1 the addition will match the floor of the existing Mall at an elevation of 228.5 feet.
Two parking structures are also part of this expansion. One of the new parking structures, the
INorth Parking Structure, will be located directly west of the present mall expansion and east of
the ring road in a paved parking area as shown on the Plan of Explorations, Figure 2, Sheet 1. It
I will have two levels and will be about 495 feet by 167 feet in plan. The grade of the lower level
of the North Parking Structure will vary,but will be approximately at an elevation of 215 feet.
I The other parking structure,the South Parking Structure, will be located in a paved parking area
west of the Sears store as shown on the Plan of Explorations, Figure 2, Sheet 2. It will have four
levels with an irregular footprint that will cover an area approximately 225 feet by 410 feet. The
Ilower level grade will be at an elevation of 223.5 feet.
1 4.0 PREVIOUS GEOTECHNICAL INVESTIGATIONS
111
Shannon &Wilson made the first of several geotechnical investigations and reports for the
e Washington Square Mall complex in January 1970. The following are geotechnical r ports that
were made by Shannon &Wilson for the Washington Square Mall.
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Preliminary Foundation Investigation, Proposed ShoppingCenter, Progress, Ore on,
February
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16, 1970, an unpublished report prepared for Winmar Company, Inc., Los Angeles,
rCalifornia.
Foundation Investigation,Progress Shopping Center, Progress, Oregon, July 10, 1970, an
1 unpublished report prepared for Winmar Company, Inc., Seattle,Washington.
Additional Foundation Investigations,Washington Square Shopping Center, Progress, Oregon,
1 September 10, 1971, an unpublished reportprepared for Winmar Company, Inc., San
p P PP
Francisco, California.
I Foundation Investigation,Washington Square Shopping Center, Progress, Oregon, February 24,
1972, an unpublished report prepared for Winmar Company, Inc. Los Angeles,
California.
I Foundation Investigation for Meier& Frank Company Retail Store,Washington Square,uare,
Progress, Oregon, March 1, 1972, an unpublished report prepared for Meier& Frank
Company, Portland, Oregon.
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Foundation Investigation for Sears Roebuck and Company Retail Complex,Washington Square,
Progress, Oregon, March 27, 1972, an unpublished report prepared for John Graham and
Company, Seattle,Washington.
Additional Subsurface Investigation,Proposed Meier & Frank Store,Washington Square,
Progress, Oregon,April 28, 1972, an unpublished report prepared for Meier& Frank
Company, Portland, Oregon.
In addition to making the geotechnical investigations, Shannon &Wilson also provided
' geotechnical observations during the grading of most of the site and the pile driving for most of
the structure foundations. All of the geotechnical reports were reviewed, and pertinent borehole
logs are included for reference in Appendix A.
5.0 FIELD EXPLORATIONS AND LABORATORY TESTING
5.1 Field Explorations
The field exploratory program consisted of 14 borings drilled at the locations shown on the Plan
of Explorations, Figure 2, Sheets 1 and 2. The locations of the borings are approximate and
based on measurements from known reference points using a measuring wheel or cloth tape.
The elevations of these borings were determined by interpolation from the topography shown on
the Plan of Explorations, Figure 2, Sheets 1 and 2.
Geo-Tech Explorations of Tualatin, Oregon drilled all of the borings using a truck-mounted
Mobile B-59 rotary drill. Borings B-1 through B-11 were drilled between April 11 and 16, 2003
and borings B-12, B-13 and B-14 were drilled on August 21, 2003. A Shannon &Wilson, Inc.
111 geologist or engineer was present throughout the explorations to collect samples and log the
borings. The borings were drilled to depths between 22.3 and 60.0 feet below the ground surface
with a combined total drilled footage of 569.8 feet.
Representative samples were obtained at approximately 2-1/2-foot to 5-foot intervals using a 2-
inch O.D. split-spoon sampler. Standard penetration testing (SPT) was performed in conjunction
with the disturbed, 2-inch O.D. split-spoon sampling in accordance with ASTM D 1586 to obtain
SPT N-values, which measure in-situ relative density and consistency. Several relatively
undisturbed samples were also obtained in the borings using a 3-inch O.D. thin walled Shelby
tube sampler in accordance with ASTM D 1587. All samples were sealed to retain moisture and
returned to our laboratory for additional examination and testing.
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Observation wells were installed in Borings B-2, B-11 and B-14 to allow the measurement of
ground water with time. Each observation well consists of a 2-inch PVC riser pipe with a 5- to
1 10-foot long slotted section at the bottom. The bottom, slotted section of the pipe is surrounded
with filter sand. The hole below the bottom section of pipe and above the sand filter was
1 backfilled with bentonite chips. A protective cover was cemented in placed at the ground
surface.
The soil classification and boring log key are described in Figure 3. Descriptive summary logs
of the subsurface borings are presented in Figures 4 through 17. Soil descriptions and interfaces
on the logs are interpretive, and actual changes may be gradual. The left-hand portion of the
boring log presents groundwater information and gives our interpretation of the soils encountered
during the field exploration program. The right-hand, graphic portion of the boring logs shows
sample locations, the results of the SPT blow counts, and sample water contents.
5.2 Laboratory Testing
• All samples returned to our laboratory were visually examined in general accordance with the
visual manual procedure (ASTM D 2488) to refine the field classifications where necessary. The
visual manual soil classification system is outlined in Figure 3. Natural water contents (ASTM
D 2216) were determined for all applicable samples, and the in-place density of the soil retained
in the Shelby tube samples (ASTM D 2937) was determined where appropriate. To assist in
evaluating liquefaction potential during a seismic event, the percent fines for selected,
representative samples was determined by washing them through a Number 200 sieve (ASTM
D1140). Atterberg Limits (ASTM D 4318) and a grain size distribution(ASTM D 422) were
rdetermined on select samples. One consolidation test (ASTM D 2435) was also performed to
assist in determining the magnitude of settlement beneath spread foundations.
The results of the water content determinations, density tests and percent fines are shown on the
boring logs, Figures 4 through 17. The results of the Atterberg limits and the grain size
distribution are shown on Figures 18 and 19, respectively. The results of the consolidation test
are shown on Figure 20.
6.0 GEOLOGY AND SEISMICITY
6.1 General Geology
Two major geologic units underlie the major portion of the Washington Square Mall. The
youngest is the Willamette Silt Formation that consists of fine-grained materials deposited by
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one or more phases of catastrophic glacial outburst floods from late Pleistocene Lake Missoula.
The Willamette Silt is generally found throughout the Tualatin Valley up to an elevation of about
250 feet, and it reaches a maximum thickness on the order of 50 feet in the center of the valley.
The Formation is composed predominately of light brown or gray-layered silts and fine sands
111 with occasional lenses of pebble and clay.
In the Washington Square area, the Willamette Silt overlies basaltic lavas of the Columbia River
Basalt Group (CRBG) that is regionally present beneath the Tualatin Valley. The CRBG
outcrops in Cooper Mountain and Bull Mountain to the south of the Washington Square Mall
complex andin the Tualatin Mountains (Portland Hills) to the north. Deep water well
information indicates that the CRBG Formation is over 500 feet thick in the vicinity of the site.
In the Tualatin Valley, lacustrine soils of the Pliocene Troutdale Formation are generally located
stratigraphically between the Willamette Silt and the CRBG. However, only two of the previous
borings made for the Sears Auto Center appear to have encountered the Troutdale Formation,
and none of the present borings encountered the Troutdale Formation.
In addition to the two major geologic units underlying the site,manmade fills are as a result of
the grading for the construction of the Mall in 1972.
6.2 Seismic Setting
Earthquakes in the Pacific Northwest occur as a result of the collision of two of the earth's great
crustal plates, the Juan de Fuca oceanic plate and the North American continental plate, and
related volcanic activity. These two plates meet along a mega thrust fault called the Cascadia
Subduction Zone. The Cascadia Subduction Zone (CSZ)runs approximately parallel to the
coastline from northernmost California to southern British Columbia. The compression forces
that exist between these two colliding plates cause the oceanic plate to descend, or subduct,
beneath the continental plate. This process leads to contortion and faulting of both crustal plates
throughout much of the western regions of Washington, Oregon, northern California, and
southern British Columbia, though the majority of the stress is relieved through great
earthquakes at the plate interface (CSZ).
6.3 Seismic Source Zones
The subduction process results in three basic types of earthquakes, each originating from
respective source zones in the earth's crust. Earthquake specialists at the Oregon Department of
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Geology and Mineral Industries have selected the following maximum plausible earthquakes for
each of the following(Mabey and others, 1993):
Shallow focus Crustal Earthquakes typically located within the upper 20 km of the continental
crust could be generated by contortion of the overriding North American plate beneath the
Portland area. Mabey and others (1993) have concluded from their analysis of local geologic
features that a crustal earthquake of up to magnitude Mw=6.5 could occur virtually anywhere in
the Portland area. We consider this magnitude to be consistent with a Maximum Credible
Earthquake (MCE) for this source zone that is typically only considered for facilities,which if
damaged, would result in catastrophic loss of life, such as nuclear power plants and large dams.
Crustal sources evaluated for this study include the potentially active Portland Hills/Clackamas
River Fault Zone.
Subduction Zone Interface Earthquakes originate along the interface between the Juan de Fuca
and North American plates as they periodically thrust by one another within the Cascadia
Subduction Zone(CSZ) located 40 km beneath the coastline. Weaver and Shedlock (1989)
selected an earthquake with a moment magnitude (Mw) of 8.5 originating 100 km from Portland
as the likely maximum plausible event from that part of the subduction zone. It is generally
believed that the subduction zone thrust fault events are relatively consistent in magnitude (MW =
8.0 - 9.0). However,paleoseismic evidence (e.g., Atwater and others, 1995) and now historic
tsunami studies (Satake and others, 1996) appear to indicate that the most recent subduction zone
thrust fault event in the year 1700 probably ruptured the full length of the CSZ and may have
approached the M =9 in size. Seismic hazard assessments should thus consider potential
earthquakes of this magnitude although evidence suggests that events of smaller size have also
occurred in the past (Atwater and others, 1995).
Deep focus, Intraplate Earthquakes originate from within the subducting Juan de Fuca oceanic
plate as a result of the downward bending and contortion of the plate. These earthquakes
typically occur 45 to 60 km beneath the surface. Such events could be as large as MW= 7.5.
However, ground shaking produced by intraplate earthquakes would be less intense and less
prolonged in the Portland area than ground motions generated by great subduction zone events as
noted above (Mabey and others, 1993), and as a result, this source zone will not be considered
afurther in this report.
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6.4 Local Seismic Sources
During middle Miocene time, flood basalt flows of the CRBG inundated northwest Oregon and
now underlie most of Multnomah, Washington and Clackamas Counties. Folding and faulting
later complexly deformed the CRBG rocks. Uplift of the CRBG formed the Tualatin Mountains
(or"Portland Hills"). The Portland Hills are elongated in a northwest southeast direction and
bounded on the northeast flank by the Portland Hills/Clackamas River Fault Zone, a right-lateral,
111 primarily strike-slip fault system.
Numerous other faults both parallel to, and transverse to, the northwest trend have been mapped
in the Portland Hills (Beeson and others, 1989, 1991). All mapped faults within the Portland
basin have cut the CRBG and were therefore active during at least middle Miocene time(about
17 million years ago). Madin (1990)reported that several faults within the Portland basin cut the
Pliocene age Sandy River Mudstone and Troutdale Formations suggesting that the faulting may
be as young as Pleistocene in age(less than 1.6 million years). Madin has most recently publicly
announced the discovery of near surface displacements in the late Pleistocene age Catastrophic
Flood Deposits.
Although there is no definitive evidence for recent activity on specific mapped faults, the
1 Portland Hills/Clackamas River Fault Zone,was judged by Geomatrix (1995) to be potentially
active on the basis of possible deformation of late Pleistocene sediments (inferred from
subsurface sediment thickness data) and the topographic expression of the Portland Hills. The
recently discovered displacements in late Pleistocene age Catastrophic Flood Deposits are
believed to be associated with the Portland Hills/Clackamas River Fault Zone and would have
occurred at sometime within the past 13,000 years.
1 6.5 Historical
Few active faults have been mapped by seismicity in Oregon,but the number is obviously
increasing with every significant seismic event. With the exception of the Cascade volcanoes,
seismicity in Oregon is generally not clearly aligned with recognized source structures
(Geomatrix, 1995).
The vast majority of recorded seismic events in Oregon are located within the crust of the North
American Plate. The Scotts Mills earthquake of March 25, 1993, with a moment magnitude
(Mw) of 5.6, is the largest instrumentally recorded earthquake located in western Oregon. At
least 17 events of MW=4 and larger have occurred in the Portland area in historic time; six events
have been estimated at Mw 5 or greater (Bott and Wong, 1993). However, the historical record
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for the Portland Region is only about 150 years long. Prior to 1980, when the University of
Washington expanded its regional network of seismographs into northwestern Oregon, few
stations operated in Oregon; and thus,very little direct information is available to determine
earthquake recurrence intervals, active fault locations, and resulting probabilistic based design
parameters.
7.0 SITE CHARACTERIZATION
7.1 Subsurface Conditions
7.1.1 General
The borings made for this investigation verified the general geologic conditions described
previously and also verified previous explorations made for the existing Mall complex. Table 1
provides a tabulation of the depth of the borings, interpreted ground elevations, fill thicknesses
where encountered, depth to rock,rock elevations, water levels where observed and date of
completion of the borings for the current and pertinent previous borings. The following sections
present our interpretation of the subsurface conditions at the new Mall Addition and at the two
parking structures.
7.1.2 Mall Addition and North Parking Structure
Nine borings, B-1 through B-6 and B-12, B-13 and B-14, were made for the new Mall
Addition and the North Parking Structure. All of these borings were made in the existing
parking area and encountered a pavement section that generally consists of 4 inches of asphalt
overlying an average of about 6 to 8 inches of granular base. However, the base thickness varied
from non-existent to 25.5 inches. Beneath the pavement sections, the borings encountered
manmade fill that was placed as a controlled, compacted fill during grading for the Mall in the
early 1970's from material cut from the higher ground on the north part of the Mall site. The fill
was found to consist of hard or dense, gravelly, clayey silt to silty clay. In boring B-14, a
boulder was encountered between a depth 5 and 6 feet. The fill was found to vary from 3 to 10
feet in thickness and appears to be thinnest at the northeast corner of the new Mall Addition and
thickest at the southwest part of the North Parking Structure.
Beneath the fill, the native soils were found to consist of loose to medium dense,
stratified sandy silt to slightly sandy, slightly plastic silt that are part of the Willamette Silt
Formation. Beneath the Willamette Silt, basalt bedrock of the CRBG was encountered at depths
varying from 17.0 to 38.0 feet. Including the previous borings, the surface of the rock varies
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from a high elevation of about 208 feet at the northeast corner of the new Mall Addition to a low
of about 186 at the southwest corner of the North Parking Structure.
Groundwater was measured in boring B-2 at a depth of 12.8 feet on April 17, 2003 and
19.0 feet on August 26, 2003. At boring B-14,the groundwater was measured at a depth of 22.4
feet on August 26, 2003. Groundwater is expected to fluctuate with the time of year,being
highest in late winter or early spring and lowest in late autumn or early winter.
7.1.3 South Parking Structure
Nine borings, B-7 through B-11, were made for the South Parking Structure. All of these
borings were made in the existing parking area and encountered a pavement section that
generally consists of 4 inches of asphalt overlying 6 to 18 inches of granular base. Beneath the
pavement sections, the borings encountered manmade fill that was placed as a controlled,
compacted fill during grading for the Mall in the early 1970's from material cut from the higher
ground on the north part of the Mall site. The fill was found to consist of hard or dense, gravelly,
clayey silt to silty clay. The fill was found to be thicker in this area than the new Mall Addition
and North Parking Structure because the original ground sloped down to the south. The fill
depth was found to vary from 19.5 to 22.5 feet in thickness.
Beneath the fill, the native soils were found to consist of loose to medium dense,
stratified sandy silt to slightly sandy, slightly plastic silt that are part of the Willamette Silt
Formation. Beneath the Willamette Silt,basalt bedrock of the CRBG was encountered at depths
varying from 38.0 feet to 54.5 feet in thickness. The surface of the rock varies from a high
elevation of about 187.0 feet at boring B-8 located at the southeast corner of the South Parking
Structure to a low of about 168.5 feet at boring B-10 located at the northwest corner of the South
Parking Structure.
Groundwater was measured in boring B-2 at a depth of 31.0 feet on April 15, 2003 and
35.4 feet on August 26, 2003. Groundwater is expected to fluctuate with the time of year, being
highest in late winter or early spring and lowest in late autumn or early winter.
7.2 Subsurface Profile
The maximum depths of the exploratory holes made at the site for this project are 55.2 feet in the
area of the new Mall Addition and North Parking Structure and 60.0 feet in the area of the South
Parking Structure. For purposes of the seismic site hazards evaluation, the subsurface is
based on the explorations made at the site, and subsurface information below this depth is based
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on published geology. Our interpretation of the subsurface profile at the new Mall Addition and
North Parking Structure site and South Parking Structure site, listed from the surface down, is as
follows:
New Mall Addition and North Parking Structure
Estimated Shear
Depth in feet Geologic Unit Wave Velocity
0 - 6 Manmade Fill 300 m/s
6 - 27 Willamette Silt 250 m/s
> 27 Columbia River Basalt 1000 m/s
South Parking Structure
Depth in feet Geologic Unit Estimated Shear
Wave Velocity
0 - 22 Manmade Fill 300 m/s
22 - 48 Willamette Silt 250 m/s
> 48 Columbia River Basalt 1000 m/s
8.0 SEISMIC HAZARDS
8.1 Design Earthquakes
The Design Earthquake referred to in the 1998 Oregon Structural Specialty Code based on the
1997 Uniform Building Code should, as a minimum,be one having a 10% probability of
' occurrence in 50 years (1 every 500 years). However, in our opinion, there is insufficient
seismic history/data to develop a probability based design earthquake, and thus a more
deterministic(subjective) method has been(and typically is) used to determine the following
design earthquakes based on the first two source zones described above. Parameters for both
design earthquakes are summarized in the following table.
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IDesign Earthquakes
Type of Mean Peak Significant Cycles Duration
I Earthquake Magnitude Distance Rock
Acceleration (Seed&Idriss 1982) (Seed&Idriss 1982)
ICrustal Mw=6.0 10 km 0.23 g 5 10 - 20 sec
CSZ M =8.5 100 km 0.12 g 26 1 - 4 min
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Using the empirical attenuation relationships developed by Sadigh, as recommended in
IGeomatrix (1995), it is estimated that local crustal faulting with a source to site distance of 10
km from the site (M„,=6.0, R=10 km)would produce mean peak rock accelerations on the order
Iof 0.23 g. The subduction zone earthquake selected (Mw 8.5, R=100 km) is estimated to result
in mean peak horizontal rock accelerations of 0.12 g in the Washington Square Mall area,based
I on the attenuation relationships proposed in Geomatrix (1995), based on data recordings of
subduction zone earthquakes.
IThe crustal source zone includes the Portland Hills/Clackamas River Fault Zone and any other
mapped or buried (or"blind") faults in the vicinity of the site. The distance of 10 kilometers was
Iselected due to the consideration of the depths at which similar magnitude faulting has taken
place, as well as the lack of active fault location data.
IGeomatrix (1995) completed a statewide"probabilistic" seismic design mapping project to be
used for new Oregon Department of Transportation structures. One of the products of this study
Iis a series of contour maps that depict mean peak bedrock accelerations. The acceleration
contours were developed by combining the contributions of all possibly active sources and
I source zones. The following mean peak bedrock accelerations, which can be considered to be
similar to ground surface accelerations, have been determined for the Washington Square Mall
i area.
Return Period Mean Peak Bedrock Acceleration
1 500 Year 0.20 g
1 1000 Year 0.28 g
2500 Year 0.39 g
I
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I8.2 Ground Response
The mean peak rock accelerations tabulated for the deterministic design earthquakes is less
Iconservative than the 1998 Oregon Structural Specialty Code Zone 3 requirement of Z=0.3 (g)
for western Oregon. In our opinion, the Structural Engineer may use either. In the event the
Engineer chooses to use the default code value of Z=0.3(g) the remaining coefficients that
account for soil amplification are recommended as follows:
ISoil Profile Type: SD
Coefficient Ca: 0.36
ICoefficient Cv: 0.54
Near Source Factor Na: 1.0
Near Source Factor Nv: 1.0
Seismic Source Type: "A" for the subduction event; "C" for the majority of crustal
I faults,though some crustal faults in Oregon may be type"B"but
are generally not yet identified (no historical quakes estimated to
be 6.5 magnitude). Nonetheless, all were considered in the
Irecommended near source coefficients recommended.
Low-rise structures (short period structures) may use the recommended design earthquake
I
acceleration (0.23g) in place of the Code Z value and be combined with the recommended
coefficients list above.
f8.3 Seismic Hazards
I8.3.1 General
We have analyzed the seismic hazards, which are specifically listed in the amended
ISection 1804.3.2 of the 1998 Oregon Structural Specialty Code. Based on the available
information, there is no reason to believe that landslides, tsunami (tidal/ocean wave) or seiche
I (oscillation of water in a lake or river) inundation are seismically induced risks to this site. The
potential hazard due to fault rupture is not known. However, the magnitude range of 6.0 to 6.5 is
the threshold at which fault rupture begins to be commonly apparent (Bonilla and others, 1984 in
IMabey and others, 1993). Because a magnitude 6.5 earthquake is the likely maximum
magnitude for any crustal earthquake in the area, fault rupture is likely to be absent or of very
I
limited extent. Liquefaction (and resulting lateral spreading and subsidence) and amplification
of earthquake shaking have been identified as potential hazards in the Washington Square area.
I
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Amplification is addressed in the preceding section. The potential for liquefaction is discussed
in the following section.
' 8.3.2 Liquefaction
Liquefaction occurs as a result of loss of strength during a seismic event in loose,
predominantly cohesionless soils such as fine sands and silts that are located below the
groundwater level. The susceptibility of a site to liquefaction is commonly expressed in terms of
a factor of safety against the occurrence of liquefaction. The factor of safety is defined as the
ratio between available soil resistance to liquefaction, expressed in terms of the cyclic stresses
required to cause liquefaction and the cyclic stresses generated by the design earthquake. The
simplified empirical procedure described by Youd (1993) was used to evaluate the susceptibility
of the site to liquefaction. For this evaluation, the two design earthquakes were considered.
Our analysis indicates that the risk of liquefaction during the magnitude 6.0 crustal
earthquake is low, but that there is a risk of liquefaction of the native soils below the
groundwater level during the magnitude 8.5 subduction zone earthquake. The new Mall
Addition and North Parking Structure will be pile supported, and as a result, loss of bearing
strength of the soil and seismically induced subsidence are not considered significant factors.
Spread foundations have been considered for the South Parking Structure,but it now appears that
it will also be pile supported. If the South Parking Structure is supported on spread foundations,
the footings may experience additional settlements due to seismically induced settlement,but
loss of bearing strength of the compacted fill supporting the spread foundations is not anticipated
because of the thick surface layer of compacted fill.
' The more significant effect of the liquefaction will be the tendency of the ground to
spread laterally toward the lower ground to the west if the zone of liquefaction is continuous. At
this time, however, it does not appear to be. This lateral spreading could result in ground
displacements at the structure that may vary from a few inches to more than a foot. More
' detailed analysis and topographic data will be required to quantify the amount of lateral
spreading. However,mitigation measures for the South Parking Structure, if supported on
' spread foundations, are discussed in a following section.
1
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I9.0 GEOTECHNICAL RECOMMENDATIONS
9.1 Earthwork
I9.1.1 General Site Preparation
I Site preparation should include the stripping and removal of all pavements,unsuitable
fill, if any, and demolition debris from areas within the footprint of the proposed structures.
I Stripping depth should be determined in the field,but for cost estimating and bidding purposes,
assume 6 inches. It may be cost effective to pulverize the existing asphalt pavement in-place and
use the resulting aggregate mix for miscellaneous structural fills, borrow for any over-
I excavations, or in place of aggregate base beneath sidewalks.
•
After stripping, the area should be graded reasonably level, and the building and
structural pavement areas should be proof-rolled or observed and probed under the observation
of the Geotechnical Engineer or his representative. Any soft or disturbed areas that are detected
ill
by the proof-rolling or probing should be removed and backfilled with engineered fill. The
actual amount of soft or disturbed material to be excavated will need to be determined in the
Ifield, and we recommend that the specifications include a unit cost bid item for any over
excavation and backfill documented by the geotechnical engineers representative.
I9.1.2 Excavations
Excavations will be required for footings or pile caps, utilities, and a portion of the North
Parking Structure, and in our opinion, these excavations can be accomplished with conventional
excavation equipment. The Contractor should be made aware of the possibility of encountering
Iscattered cobbles and boulders as found in boring B-14. Because of safety considerations and
the nature of temporary excavations,the Contractor should be made responsible for maintaining
Isafe temporary cut slopes and supports for utility trenches, etc. We recommend that the
Contractor incorporate all pertinent safety codes during construction. Due to the shallow depth
Iof the excavations, it is likely that safe slopes may be preferred over shoring.
9.1.3 Temporary Dewatering
IWe recommend that the Contractor check the groundwater level prior to excavation. If
I dewatering is found to be necessary, it should be made the Contractor's responsibility to provide
an acceptable means of dewatering. The dewatering should be accomplished in a manner that
will not adversely affect excavation stability and subgrade integrity. In our opinion, pumping
Ifrom open sumps may be possible in the more cohesive soils or where the depth of water to be
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Ilowered is small. However, should the foundation soils become "quick," a more sophisticated
dewatering system such as well points or wells will be required. The Owner's representative
I should be given the opportunity to review and comment on any dewater methods proposed by
the Contractor.
1 9.1.4 Structural Fill
It does not appear that significant fill will be required to raise the site grade, however,
Iwhere fill is required, any soil, including the on-site soil that is free of organic or other
deleterious matter will be suitable provided it is placed during dry warm weather and it is
Imoisture conditioned, if necessary, (to raise or lower the water content)before it is placed to
achieve optimal moisture content for compaction. For fill under footings, we recommend that a
I granular fill be used. If grading work is accomplished during the wet time of the year, then a
clean(not more than 7 percent passing the No. 200 sieve, based on a wet sieve analysis),
reasonably well-graded, granular soil is recommended. The on-site soils, in our opinion, are not
acceptable as fill during wet weather, and we recommend, if at all possible, that earthwork be
scheduled for the dry summer or fall months of the year.
I
General fills should be placed in thin lifts and compacted to a dry density of at least 95
I percent of the standard Proctor(ASTM D 698)maximum dry density. Structural fills beneath
footings or within a 2-foot depth of any traveled pavement sections should be compacted to 98
I percent of the standard Proctor maximum dry density. Landscape fills should be compacted to a
minimum of 90 percent of the standard Proctor maximum dry density. The thickness of the lifts
will need to be determined in the field, but generally for larger self-propelled compactors, the
Ilifts should not exceed about 9 to 12 inches as measured in a loose condition. For small plate
compactors, the lifts should not exceed 4 inches loose measure.
1 9.1.5 Slopes
I All permanent cut and fill slopes should be graded to 1 vertical on 2 horizontal or flatter.
Flatter slopes maybe necessary for ground cover and maintenance operations.
I9.2 Foundation Design
9.2.1 General
IIt is our understanding that the new Mall Addition and North Parking Structure will be
I supported on driven piles. Spread foundations have been considered for the South Parking
Structure,but it now appears that it will also be pile supported. The Mall Addition will be a one
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1 and two level structure with maximum column loads of 250 kips and maximum wall loads of 5
kips per lineal foot. The North Parking Structure will have two levels with maximum column
loads of 300 kips and maximum wall loads of 7 kips per lineal foot. The South Parking Structure
will be a four level building with three framed levels. Assuming a column bay spacing of 60 feet
by 30 feet, the maximum center column loads are 750 kips, maximum edge column loads are 400
kips, and maximum wall loads are 20 kips per lineal foot. It is anticipated that if the South
Parking Structure is supported on spread foundations, they will be strip footings running the
length of the structure.
' 9.2.2 Pile Foundations
9.2.2.1 Pile Types
111 The type of pile and allowable load capacity are typically dependent on design
loads, soil conditions and economic considerations. The existing Mall complex is supported on
Idriven 10-inch nominal diameter steel pipe piles. Other types of piles such as steel H-piles and
pre-cast concrete piles or driven, cast-in-place grout piles are also considered suitable. The pre-
cast concrete pile, however, may not be economical because of the variable pile lengths that will
be required. In addition, we do not recommend augured, cast-in-place piles or treated timber
piles. The augured, cast-in-place piles may not penetrate the basalt rock a sufficient distance to
develop design bearing, and the timber piles may be prone to damage. For ease of installation,
we recommend either a thick walled steel pipe pile driven closed ended with a steel plate welded
on the bottom or a steel H-pile with suitable tip protection. These two pile types can be easily
cut or spliced in the field. Based on correspondence with DLR Group, we understand that the
piles will be steel pipe piles with a nominal diameter of 10 inches and a wall thickness of 318th
inch. Therefore, the following paragraphs regarding pile foundations assume that 10-inch
nominal size steel pipe piles will be used.
9.2.2.2 Pile Capacity and Driving Criteria
The piles will be relatively short and will be driven to end bearing in the basalt
bedrock. Therefore, the allowable pile capacity can be in the range of 50 to 100 tons or greater,
and the size of the pile will depend on the allowable pile capacity selected for design. The 1998
Oregon Structural Specialty Code requires that the allowable axial stresses shall not exceed 0.35
' of the minimum specified yield strength,Fy, or 12,600 pounds per square inch (psi), whichever is
less. Assuming that the 12,600 psi criterion governs, the allowable load for a 10-inch nominal
size steel pipe piles is 75 tons.
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We recommend that the end bearing piles be driven to a driving resistance
I
determined by the pile driving criteria developed by the Oregon Department of Transportation
I (ODOT), Section 0052, Standard Specifications for Highway Construction. Pile driving
equipment and a pile driving formula are part of this specification. We recommend that the
Geotechnical Engineer of Record be retained to observe the driving of the piles and provide
Iquality control.
9.2.2.3 Estimated Pile Lengths
1
Since the 10-inch nominal size pipe piles will be driven basically to refusal in the
Ibasalt bedrock, the pile lengths will be the difference between the bottom of the pile cap and top
of the bedrock plus whatever distance the pile penetrates into the rock. The depth of penetration
of the pile into bedrock depends on the degree the surface of the rock is weathered. Based on our
experience in the area, we estimate that the piles will penetrate 2 to 5 feet into the rock.
Assuming that the bottom of the pile cap is 3 feet lower than the finished floor of the structures
and an average penetration of 3 feet, it is estimated that the piles for the Mall Addition will vary
from about 18 to 42 feet in length to cutoff. At the North Parking Structure, it is estimated that
Ithe piles will vary from about 10 to 33 feet in length to cutoff, and at the South Parking
Structure, it is estimated that the piles will vary from about 36 to 55 feet in length to cutoff.
I9.2.2.4 Uplift Capacity of Piles
I We understand that the uplift capacity of the piles is required for the South
Parking Structure. The uplift capacity of the piles is derived from the adhesion between the steel
pile and the upper clayey fill and the friction between the steel pile and the native sandy silts. As
I a result, the uplift capacity of the piles will be dependant on the driven length of the piles. At the
South Parking Structure, it is anticipated that the piles will vary in length from about 36 to 55
Ifeet. Allowable uplift capacity resulting from the adhesion and friction of the soil will range
from 22 kips for a 30-foot long pile to 54 kips for a 60-fool long pile. Although the relationship
Ibetween uplift capacity and pile length is not linear, it is close enough to assume that it is. In
addition to the adhesion and friction from the soil, the weight of the pile may be included as a
resisting force.
I
If the uplift resistance of the piles is not sufficient, rock anchors grouted into the
Ibasalt bedrock is another option.
I
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1 9.2.2.5 Lateral Pile Capacities
Lateral load capacities for the 10-inch nominal size pipe piles were determined
Iusing the simplified procedures provided in NAVFAC DM-7.2. The lateral load on a single pile
varies with the pile length. Very short piles have little or no lateral resistance since they will
I tend to tilt rather than bend. Our analysis indicates that piles shorter than about 20 feet will have
limited lateral load capacity if they are driven into the same soil unit. However, the piles will be
I driven into the weathered portion of the basalt bedrock a few feet, and as a result should be
restrained from tilting. Therefore, we do not believe a reduction will be necessary for the shorter
piles.
I
The lateral loads for the piles also depend on the deflection allowed at the top of
the pile. The larger the allowable deflection, the greater the lateral capacity. For our analysis,
we have determined the lateral capacity for a single, 10-inch nominal size pipe pile for an
I allowable defection of 0.25 and 0.5 inches. For an allowable deflection of 0.25 inches, the
lateral capacity of the pile is 8.7 kips, and for an allowable deflection of 0.5 inches,the lateral
capacity of the pile is 17.4 kips.
I
The lateral loads given in the previous paragraph are for a single pile. Group
action needs to be considered when the pile spacing in the direction of loading is less that 6 to 8
I pile diameters. For pile groups, we recommend a minimum center to center pile spacing of 3
pile diameters. The first pile in the direction of loading carries the full lateral capacity. Each
Isubsequent pile at a pile spacing of 3 pile diameters behind the first pile in the direction of
loading will have a reduced lateral capacity. For an allowable deflection of 0.25 inches, the
Ireduced lateral capacity is 3.5 kips per pile, and for an allowable deflection of 0.5 inches, the
reduced lateral capacity is 7.0 kips per pile. Additional lateral load capacity may be developed
Ithrough passive pressure against the pile cap. The ultimate passive resistance may be computed
using an equivalent fluid pressure of 250 pounds per cubic foot.
1 9.2.3 Spread Foundations
If shallow foundations are used for the South Parking Structure, we understand that a mat
foundation is not possible due to the 60-foot column spacing, but continuous strip footings could
be designed at 60-foot centers. For spread foundations bearing on the fill that was previously
Iplaced over the site area, we recommend that the allowable bearing pressure should not exceed 3
kips per square foot (ksf), and when sizing footings for seismic considerations, the allowable
Ibearing pressure may be increased by one-third. Where the fill is absent because it is removed
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SHANNON FiWILSON.INC.
during site grading, the allowable bearing pressure should not exceed 2 ksf. For design of the
strip footings, we also recommend a modulus of subgrade reaction of 150 pounds per square inch
per inch.
Continuous wall footings should have a minimum width of 18 inches, and column
footings should have a minimum width of 24 inches. All perimeter footings should be founded
at least 24 inches below the lowest adjacent grade, which should be taken as the lower of
finished floor elevation or exterior grade. Interior footings may be founded at a depth of 18
inches below finished floor elevation.
In order to mitigate the potential for lateral spreading during the magnitude 8.5
subduction zone earthquake if spread foundations are used for the South Parking Structure, we
1 recommend that the continuous strip footings be tied together with grade beams.
We estimate that total settlement of the strip footings for the South Parking Structure will
1 be on the order of 2.5 inches. For spread foundations for other appurtenant structures designed
for an allowable bearing pressure of 2 ksf,we estimate a total settlement of 1-inch or less. The
settlement will be greatest at the center of the strip footing and will be about a half of the
maximum at the ends. Differential settlement between the strip footings and/or spread footings
' will be about one-half to three-quarters of the maximum.
Each footing excavation should be evaluated by a qualified Geotechnical Engineer to
confirm suitable bearing conditions and to determine that all loose materials, organics, unsuitable
fill and softened subgrade, if present,have been removed. If miscellaneous fills or unsuitable
' soils are encountered at footing locations, we recommend that the fill or unsuitable soil be
removed. In order to keep footings at a conventional grade, the excavated fill beneath footings
may be replaced with compacted structural fill to grade as shown by Figure 21. To minimize the
potential for disturbance during excavation for the footings, we recommend that excavations be
made with a smooth bucket (no teeth)backhoe.
9.2.4 Floor Slabs
All floor slabs on grade, should be founded on a minimum 6-inch layer of free-draining,
well-graded sand and gravel or crushed rock with a maximum particle size of 1-1/2 inches and
containing not more than 4 percent passing the No. 200 sieve (based on a wet sieve analysis).
The base rock is to be compacted to a dry density of at least 95 percent of the standard Proctor
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Imaximum dry density (ASTM D 698). A vapor barrier is recommended for all heated, finished
areas.
IWe understand that the existing Mall has a system of underslab drains that were installed
to relieve potential hydrostatic head resulting from a combination of high groundwater and
Isubstantial cuts over the north leg of the existing L-shaped Mall. The February 24, 1972
geotechnical report recommends underslab drains for the Mall and presents a possible layout that
I only shows the north leg of the L-shaped Mall. A review of the original grading plan also shows
that most of the current site is located over an area that was filled, and the cut that was made is
I only a few feet at most. As a result of our findings, we do not believe that underslab drains are
necessary for the new Mall Addition.
1 9.3 Embedded Walls
Embedded walls such as small cantilever retaining walls or basement walls should be designed to
Iresist lateral pressures. The lateral pressure will depend on the ability of the walls to yield.
Conventional retaining walls should be designed for an equivalent fluid weighing 35 pounds per
Icubic foot (pcf). Non-yielding walls should be designed for an equivalent fluid weighing 50
pounds per cubic foot (pcf). These values assume that the wall is properly drained to prevent the
I buildup of hydrostatic pressures and that the back slope is horizontal. Higher lateral pressures
may be anticipated for up-slope backfill.
1 Sliding, overturning, and maximum toe pressure should be checked for cantilever walls. The
lateral pressure may be resisted in part from the passive resistance of the soil in front of the wall
I footing. Ultimate passive resistance may be computed on the basis of 250 pounds per cubic foot
equivalent fluid where horizontal ground surfaces prevail, provided that the ground surface in
I
front of the footing is thoroughly compacted. If there is a possibility that future excavations
could occur in front of the wall for the installation of utilities, the top 2 to 3 feet of the passive
resistance should be disregarded. Additional ultimate resistance to lateral earth pressure may be
Iobtained from sliding resistance of the base of the wall footing. We recommend a friction factor
of 0.35 for the silt subgrade to determine the sliding resistance at the base. We recommend that a
1 factor of safety of 1.5 be used for sliding and overturning.
I Drainage is considered necessary to protect against saturation of the backfill due to leakage from
broken water or sewer lines. Recommendations concerning backfill and drainage requirements
behind basement and small cantilever retaining walls are shown in Figure 22. The perimeter
Idrain lines should be adequately sloped to allow the water to drain under gravity or a pump
111
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presence or absence of hazardous or toxic materials in the soil, surface water, groundwater, or
air, on or below the site, or for evaluation of disposal of contaminated soils or groundwater,
should any be encountered, except as noted in this report.
Shannon & Wilson, Inc. has prepared a document, "Important Information About Your
Geotechnical/Environmental Report,"to assist you and others in understanding the use and
limitations of our report. This document is included at the end of this report in Appendix B.
SHANNON & WILSON, INC.
PROper
%/0, �e'NEe
72842PE
OREGO
EXPIRES: 7-AVO_At
Daniel E. Hogan, P.E.
Geotechnical Engineer IV
5382
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Ecp,«s /z/3f/G 3
K. Frank Fujitani, P.E.
Vice President
11
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